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Tenth U.S. National Conference on Earthquake EngineeringFrontiers of Earthquake EngineeringJuly 21-25, 2014

Anchorage, Alaska10NCEE

A MODEL FOR PREDICTING PANEL ZONE

DEFORMATION CAPACITY INREHABILITATED STEEL MOMENT

CONNECTIONS

Dong-Won Kim1, Chia-Ming Uang

2, and Colin Blaney

3

ABSTRACT

Three full-scale specimens that simulated rehabilitated pre-Northridge steel moment connections

were tested to failure. A Kaiser Bolted Bracket (KBB) was used on the beam bottom flange for

all specimens, but different rehabilitation schemes [KBB, complete-joint-penetration (CJP)

replacement weld, or welded double-tee bracket with CJP replacement weld] were used for thebeam top flange. Test results showed that the proposed rehabilitation schemes adequately

protected the pre-Northridge moment connections to an acceptable story drift angle. Large panel

zone deformation with significant yielding occurred in all specimens. An analytical model wasdeveloped to predict the panel zone deformation capacity and the associated strength. In the

proposed model, it was postulated that the notch-tough CJP welds located at the column kinking

locations would fracture when the column flange was fully yielded there; this limit state wasused to define the ultimate deformation capacity of the panel zone. The proposed model not only

correlated very well with the test results but also showed that the deformation capacity is a

function of db/tcf, where dbis the beam depth, and tcfis the column flange thickness. The effect of

column axial load on the panel zone shear deformation was also considered in the model.

1Ph.D Candidate, Dept. of Structural Engineering, University of California, San Diego, La Jolla, CA 920922Professor, Dept. of Structural Engineering, University of California, San Diego, La Jolla, CA 920923Executive Principal, ZFA Structural Engineers, San Francisco, CA 94104

Kim DW, Blaney C, Uang, CM. A Model for Predicting Panel Zone Deformation Capacity in Rehabilitated Steel

Moment Connections. Proceedings of the 10thNational Conference on Earthquake Engineering, Earthquake

Engineering Research Institute, Anchorage, AK, 2014.

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A Model for Predicting Panel Zone Deformation Capacity in

Rehabilitated Steel Moment Connections

Dong-Won Kim1, Chia-Ming Uang2,and Colin Blaney3

ABSTRACT

Three full-scale specimens that simulated rehabilitated pre-Northridge steel moment connections

were tested to failure. A Kaiser Bolted Bracket (KBB) was used on the beam bottom flange for allspecimens, but different rehabilitation schemes [KBB, complete-joint-penetration (CJP)

replacement weld, or welded double-tee bracket with CJP replacement weld] were used for the

beam top flange. Test results showed that the proposed rehabilitation schemes adequatelyprotected the pre-Northridge moment connections to an acceptable story drift angle. Large panel

zone deformation with significant yielding occurred in all specimens. An analytical model was

developed to predict the panel zone deformation capacity and the associated strength. In theproposed model, it was postulated that the notch-tough CJP welds located at the column kinking

locations would fracture when the column flange was fully yielded there; this limit state was used

to define the ultimate deformation capacity of the panel zone. The proposed model not only

correlated very well with the test results but also showed that the deformation capacity is a

function of db/tcf, where dbis the beam depth, and tcfis the column flange thickness. The effect ofcolumn axial load on the panel zone shear deformation was also considered in the model.

Introduction

Rehabilitation of pre-Northridge steel moment connections to avoid brittle fracture in the

beam flange weld joints is a challenging task to design engineers. Designed before Pre-Northridge earthquake, it is also not uncommon that very weak panel zone exists in these

connections.

To support the seismic rehabilitation of a steel moment frame building constructed before

1994, three full-scale specimens were tested. Test results showed that panel zone yielding wasthe primary source of energy dissipation. Two specimens eventually experienced fracture at the

notch-tough CJP welds at the column kink locations due to excessive panel zone deformation.This motivated the development of a procedure to predict the ultimate panel zone deformation

capacity beyond which beam flange CJP weld fracture is imminent.

1Ph.D Candidate, Dept. of Structural Engineering, University of California, San Diego, La Jolla, CA 920922Professor, Dept. of Structural Engineering, University of California, San Diego, La Jolla, CA 920923Executive Principal, ZFA Structural Engineers, San Francisco, CA 94104

Kim DW, Blaney C, Uang, CM. A Model for Predicting Panel Zone Deformation Capacity in Rehabilitated Steel

Moment Connections. Proceedings of the 10thNational Conference on Earthquake Engineering, Earthquake

Engineering Research Institute, Anchorage, AK, 2014.

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Test Program

Test Specimens

Three nominally identical pre-Northridge moment connections with a W36150 beam

and a W14193 column were rehabilitated and tested. As a rehabilitation scheme, a KaiserBolted Bracket (KBB) was used on the beam bottom flange for all specimens, but differentrehabilitation schemes were used to strengthen the beam top flange; this included the use of

another KBB for Specimen 1, a notch-tough complete-joint-penetration (CJP) beam flange

replacement groove weld for Specimen 2, as well as a welded double-tee bracket and areplacement weld for Specimen 3 [4, 9]. Table 1 shows the mechanical properties of the

materials obtained from tensile coupon tests for the specimens.

Table 1. Steel Mechanical Properties

ComponentYield Stress

(MPa)

Tensile Strength

(MPa)

Elongation*

(%)

Beam Flange (W36150) 425 523 33

Column Flange (W14193) 428 555 31.5

* based on a 51-mm gage length

Fig. 1 shows the test setup. A corbel was bolted to the end of the beam and attached totwo 980-kN hydraulic actuators. With some minor modification, the loading sequence specified

in Appendix S6.2 of AISC 341-10 [1] for beam-to-column moment connection test was used.

Since the target story drift for this rehabilitation project was 3.5% story drift, the AISC loadingprotocol was modified to include two additional cycles at 3.5% story drift.

Column

4,420

2,286

2,286

Beam

(W14x193)

(W36x150)

Corbel

Two 980-kNHydraulic

Actuators

Units: mm

Bracing

LocationA

Bracing

Location B

Figure 1. Test Setup (Specimen 3)

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Test Results

Significant shear yielding in the panel zone was observed in all three specimens. The

KBBs remained intact and showed no sign of yielding or damage. For Specimen 1, the doubleKBBs forced beam plastic hinging in the form of flange and web local buckling as well as

lateral-torsional buckling near the tip of the KBBs.

Specimen 2 experienced significant yielding in the panel zone, but the extent of beam

plastic hinging was very limited with no sign of buckling. After completing one cycle at 4%story drift, fracture of beam top flange at the replacement weld occurred during the second cycle.

The behavior of Specimen 3 was similar to that of Specimen 2, i.e., inelastic action occurredmainly in the panel zone. The CJP weld connecting the horizontal plate of the double-tee bracket

to the column flange started to fracture at 4% story drift. Brittle fracture occurred during the

cycle of 4.5% story drift (see Fig. 2).

-300 -200 -100 0 100 200 300-1500

-1000

-500

0

500

1000

1500

Beam Tip Displacement (mm)

AppliedLoad

(kN)

-6 -4 -2 0 2 4 6Story Drift Ratio (%)

Fracture

(a)

-300 -200 -100 0 100 200 300-1500

-1000

-500

0

500

1000

1500

Beam Tip Displacement (mm)

AppliedLoad(kN)

-6 -4 -2 0 2 4 6Story Drift Ratio (%)

Fracture

(b)

Figure 2. Test Results: (a) Specimen 2; (b) Specimen 3

Effective Depth of Extended Panel Zone in Rehabilitated Steel Moment Connection

Large panel zone deformation caused the column flange to kink at four corners of the

panel zone. With the addition of KBBs and double-tee bracket, the panel zone was extended in

depth. AISC 358 [2] defines the effective depth, effd , of the extended panel zone as the centroid

distance between column bolt groups in the KBBs. Generalizing the AISC definition to

Specimens 2 and 3, the definition of effd is shown in the Fig. 3.

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CJP Weld and Panel

Zone Kink Location

deff

CJP Weld and Panel

Zone Kink Location

deff

(a) (b)

Figure 3. Effective Depth of Extended Panel Zone: (a) Specimen 2; (b) Specimen 3

Normalized Shear Deformation,y

-0.04 -0.02 0.0 0.01 0.03-4000

-3000

-2000

-1000

0

1000

2000

3000

4000

Panel Zone Shear Deformation (rad.)

ShearForce(kN)

-10 -5 0 5 10

Normalized Shear Deformation, y

-0.04 -0.02 0.0 0.01 0.03-4000

-3000

-2000

-1000

0

1000

2000

3000

4000

Panel Zone Shear Deformation (rad.)

ShearForce(kN)

-10 -5 0 5 10

(a) (b)

Figure 4. Cyclic Response of Extended Panel Zone: (a) Specimen 2; (b) Specimen 3

The shear in the extended panel zone can be computed as follows:

c

eff

fV

d

MV

95.0 (1)

whereMf= moment at the face of column, and Vc= shear in the column. The cyclic responses of

the extended panel zones for Specimens 2 and 3 are presented in Fig. 4.

Panel Zone Shear Deformation Capacity

Panel zone behavior was extensively researched [5, 6, 7, 8, 10, and 11]. However, past

research was mainly focused on the strength, not the ultimate deformation capacity beyondwhich excessive kinking in the column flanges would cause fracture of the beam flange CJP

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welds.

It was obvious from Fig. 4 that a panel zone could deform to a deformation level much

higher than 4y, a deformation level corresponding to the panel zone strength in AISC 360 [3].

But excessive deformation could cause fracture in the beam flange-to-column flange CJP weld.

To predict the panel zone deformation and the associated strength, an alternative model isdeveloped. The panel zone behavior is established by superimposing the responses of the column

web and flanges (see Fig. 5). The web area is taken as 0.95dctcw. Therefore, the shear yield

strength of the column web is

cwcyycw tdFV 95.06.0, (2)

With GFyy /6.0 , the elastic shear stiffness of the column web is:

GtdV

K cwcy

ycw

cw 95.0,

(3)

Since strain hardening generally exists for the steel grades ( 345yF MPa) permitted in

AISC 341-10 [1], a shear strain hardening ratio of 0.03 is considered as shown in Fig. 5(a). The

strain hardening ratio of 0.03 is based on monotonic torsional coupon test results conducted by

Slutter [12].

Since each column flange in the panel zone region would bend about its weak axis in

reverse curvature, the model in Fig. 6(a) is used to consider the contribution from columnflanges. It is idealized that each column flange will deform elastically until the plastic moment of

the column flange is reached:

yc

cfcf

cfp Ftb

M

4

2

, (4)

The associated panel zone deformation, which is the chord angle in Fig. 6(b), corresponds

to pzin Figure 5. It is postulated that pzcan be defined as the plastic deformation capacity of the

panel zone beyond which the notch-tough CJP weld at the kink locations is prone to fracture.This postulation is to be verified by test data. The derivation of pzfollows.

+ =

cwV

y ypz pz pzy

pzV

yV

cwV

ycw

V,

cfV

cfpV ,

cfK2

GtdK cwccw 95.0

pzV

cwK03.0

(a) (b) (c)

Figure 5. Superposition of Proposed Shear Strength Components: (a) Column Web Panel Zone

Response; (b) Response of Two Column Flanges; (c) Superposition

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b

cfpcfp

d

MV

95.0

2 ,,

cfpM ,

Moment Diagram

2

95.0b

d

cd95.0

bd95.0

pz pz

cfpM ,

(a)

Figure 6. Panel Zone Model: (a) Panel Zone Deformation; (b) Mathematical Model

Consider one fix-ended column-flange flexural member with a span of 0.95db and a

depth oftcf(see Fig. 6). The shearing effect of this flexural member can be significant when thespan (= 0.95db) is small and the column flange (tcf) is thick. Applying elastic beam theory, the

mid-span deflection when the fixed-end moment reachesMp,cfis:

cfcf

cfp

cfcf

cfp

cf

b

b

b

cfp

cfs

b

b

cfp

cf

tEb

M

tEb

M

t

d

d

d

M

GA

d

d

M

EI

,2,

2

2

,

,

3

,

12.311.1

12.311.1

2

95.0

2/95.0

1

2

95.0

2/95.03

1

(5)

where Icf (= 12/3

cfcftb ) and cfsA , (= 6/5 cfcftb ) are the moment of inertia and shear area of one

column flange, respectively. In the above equation, the coefficient is the related span-depthratio of the column-flange flexural member:

cfb td / (6)

The first term on the right-hand side in Eq. 5 is the flexural component and the second term is the

shearing component. Dividing by 2/95.0 bd and simplifying gives the ultimate deformationcapacity of the panel zone (see Fig. 6):

45.3475.0

E

Fycpz (7)

The elastic stiffness of one column flange is

45.3

11.1

95.0

22

,,

cfcf

pzb

cfp

pz

cfp

cf

tEb

d

MVK (8)

The total elastic stiffness for both column flanges is cfK2 , as shown in Fig. 5(b). Therefore, the

(b)

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total panel zone shear strength in the elastic range is

2 cfcwpz KKV when y0 (9)

When pzy , the components of panel zone shear strength due to column web and two

column flanges are [see Fig. 5(a)]

ycwycwcw KVV 03.0

,

(10)

cfcf KV 2 (11)

Therefore, the total panel zone shear strength is

cfcwpz VVV when pzy (12)

0 2 4 6 8 10 120

500

1000

1500

2000

2500

3000

3500

4000

Sh

earForce(kN)

Normalized Shear Deformation, y

pz/y

(= 8.99)

TestProposedAISC

0 2 4 6 8 10 120

500

1000

1500

2000

2500

3000

3500

4000

ShearForce

(kN)

pz/y

(= 10.04)

Normalized Shear Deformation, y

TestProposedAISC

(a) (b)

Figure 7. Comparison of panel zone responses: (a) Specimen 2; (b) Specimen 3

Based on Eqs. 9 and 12 and replacing bd with effd , the predicted panel zone responses

for Specimens 2 and 3 up to pz are shown in Fig. 7. The predicted panel zone behaviors

reasonably match well. The ratios between the predicted and experimental panel zone ultimate

deformations are 1.02 and 0.94 for Specimens 2 and 3, respectively.

Normalizing the panel zone ultimate deformation capacity, pz , in Eq. 7 by y (=

0.6Fy/G) gives the following:

45.330.0

y

pz (13)

Fig. 8 shows the variation of the normalized panel zone ultimate deformation with respect to (= db/tcf). It is shown that the AISC assumed panel zone deformation capacity (4y) can be veryconservative for a high db/tcfratio. When the db/tcfratio is low (i.e., a shallow beam connected to

a thick column flange), on the other hand, the panel zone deformation can be lower than 4y.

Therefore, column flanges at kink locations would yield early when db/tcf is low, which makesthe beam flange-to-column flange CJP welds more prone to fracture at a low panel zone

deformation ( y 4 ).

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0 10 20 30 40 500

5

10

15

20

Normalized

PanelZone

Deformation,

pz

/y

AISC (4y)

(= db/tcf)

Eq. (13)

Figure 8. Relationship between Panel Zone Ultimate Deformation Capacity and

0 10 20 30 40 500.0

0.01

0.02

0.03

0.04

0.05

Column Flange Span-Depth Ratio, (= db/tcf)

PanelZoneDeformation,pz

P/2Py,cf= 0P/2Py,cf= 0.25

P/2Py,cf= 0.5

P/2Py,cf= 0.75

P/2Py,cf= 0P/2Py,cf= 0.25

P/2Py,cf= 0.5

P/2Py,cf= 0.75

Figure 9. Effect of Column Axial Load on Panel Zone Shear Deformation Capacity (A992 Steel)

Effect of Column Axial Force

With the presence of an axial load, Krawinkler et al. [7] reported that column flanges

carry all the axial load after the panel zone web has completely yielded. This is also the basis of

the panel zone design shear strength with high axial load in AISC 360-10 [2]. Based on the sameassumption, one can derive the reduced moment capacity of one column flange as

2

,

,,

2

1cfy

cfpcfp

P

PMM (14)

Therefore, the corresponding shear of one column-flange flexural member in Fig. 6 is

2

,

,

,

,2

195.0

2

cfy

cfp

b

cfp

cfpP

PV

d

MV (15)

Following the similar procedure described in Eqs. 5 and 7, the reduced plastic shear deformation

can be derived by replacing cfpM , and cfpV , by cfpM , and cfpV , :

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2

,

2

, 21

21

45.3475.0

cfy

pz

cfy

yc

pzP

P

P

P

E

F (16)

Fig. 9 shows the effect of column axial load on the panel zone ultimate deformation capacity.

The associated panel zone shear strength at pz is established as follows. The componentof panel zone shear strength due to column web from Eq. 10 can be approximated as

ypzcwycwcw KVV 03.0, (17)

From Eq. 15, the component of the panel zone shear strength due to two column flanges is

2

,

,,2

122cfy

cfpcfpcfP

PVVV (18)

Therefore, the total panel zone shear strength is

cfcwpz VVV (19)

Fig. 10 shows example plots of the panel zone axial-shear interaction curves. A

W36150 beam with three different W14 column sections in Fig. 10(a) and W36 column

sections in Fig. 10(b) are considered. It is observed that axial load has a significant effect on thepanel zone deformation capacity than on the shear strength. Since the interaction between axial

load and panel zone shear strength is relatively weak, the axial load effect can be ignored for

simplicity when 6.02/ , cfyPP (or ,/ 1.2y cfP P )..

0.0 0.2 0.4 0.6 0.8 1.00.0

0.2

0.4

0.6

0.8

1.0

W14X109 ColumnW14X193 ColumnW14X370 Column

W14X109 ColumnW14X193 ColumnW14X370 Column

Normalized Axial Load, P/2Py,cf

NormalizedPan

elZone

ShearStrength,

V p

z

/Vpz

0.0 0.2 0.4 0.6 0.8 1.00.0

0.2

0.4

0.6

0.8

1.0

W36X210 ColumnW36X231 ColumnW36X361 Column

W36X210 ColumnW36X231 ColumnW36X361 Column

Normalized Axial Load, P/2Py,cf

NormalizedPanelZone

ShearStrength,

V p

z

/Vpz

(a) (b)

Figure 10. Interaction of Shear and Axial Force: (a) W14 Columns; (b) W36 Columns

Summary and Conclusions

An analytical model was developed to predict the panel zone ultimate deformation

capacity and the associated strength in rehabilitated steel moment connections; the ultimate

deformation was defined as that beyond which excessive kinking in the column flanges wouldcause weld fracture in the beam CJP welds. The effect of column axial load was also included in

the formulation. The conclusions are summarized as follows.

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(1) The proposed model (see Eqs. 7 and 13) showed that the ultimate deformation

capacity is a function ofdb/tcf. The low db/tcfratio (i.e., a shallow beam connected to a columnwith thick flanges) may result in earlier yielding of the column flanges at the kink location and

makes the CJP welds more vulnerable to fracture.

(2) The column axial load effect on the panel zone ultimate deformation capacity canbe significant (see Eq. 16). But the effect on shear strength is relatively insignificant (see Fig. 10)and can be ignored when the column axial load is less than 1.2 times the yield force of one

column flange.

Acknowledgements

This project was sponsored by Civic Facilities Division of the City of Fremont.

References

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2. AISC, Prequalified Connections for Special and Intermediate Steel Moment Frames for Seismic Applications,ANSI/AISC 358-10, American Institute of Steel Construction, IL, 2010.

3. AISC, Specification for Structural Steel Buildings,ANSI/AISC 360-10, American Institute of Steel Construction,IL, 2010.

4. Blaney, C., Uang, C.M., Kim, D.W., Sim, H.B., and Adan, S.M. Cyclic Testing and Analysis of Retrofitted Pre-Northridge Steel Moment Connections Using Bolted Brackes Proceedings,SEAOC Annual Convention, 2010.

Sacramento, CA.

5. El-Tawil, S., Vidarsson E., Mikesell T., and Kunnath, S.K. Inelastic Behavior and Design of Steel Panel Zones.

J. Struct. Div., ASCE1999; 125(2): 183-193.

6. Kato, B., Chen, W.F., and Nakao, M. Effects of Joint-panel Shear Deformation on Frames.J. Construct. SteelResearch1998; 10: 269-320.

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8. Krawinkler, H. Shear in Beam-Column Joints in Seismic Design of Steel Frames.Engineering Journal1978; 5(3): 82-91, American Institute of Steel Construction, IL.

9. Kim, D.W., Sim, H.B., and Uang, C.M. Cyclic Testing of Steel Moment Connections for Seismic Rehabilitationof Fremont Police StationReport No. TR-10/012011. Dept. of Struct. Engrg., Univ. of Calif., San Diego, CA.

10. Lee, D., Cotton, S.C., Hajjar, J.F., Dexter, R.J., and Ye, Y. Cyclic Behavior of Steel Moment-ResistingConnections Reinforced by Alternative Column Stiffener Details II. Panel Zone Behavior and Doubler Plate

Detailing.Engineering Journal2005; 42(4): 215-238, American Institute of Steel Construction, IL.

11. Schneider, S.P., and Amidi, A. Seismic Behavior of Steel Frames with Deformable Panel Zones.J. Struct. Div.,ASCE1998; 124(1): 35-42.

12. Slutter, R.G. Test of Panel Zone Behavior in Beam-Column Connections,Report No. 200.81.403.11981, FritzEngineering Laboratory, Lehigh University, Bethlehem, PA.